土木工程建筑外文文献及翻译
Cyclic behavior of steel moment frame connections under varying axial load and lateral displacements
Abstract
This paper discusses the cyclic behavior of four steel moment connections tested under variable axial load and lateral displacements. The beam specim- ens consisted of a reducedbeam section, wing plates and longitudinal stiffeners. The test specimens were subjected to varying axial forces and lateral displace- ments to simulate the effects on beams in a Coupled-Girder Moment-Resisting Framing system under lateral loading. The test results showed that the specim- ens responded in a ductile manner since the plastic rotations exceeded 0.03 rad without significant drop in the lateral capacity. The presence of the longitudin- al stiffener assisted in transferring the axial forces and delayed the formation of web local buckling.
1. Introduction
Aimed at evaluating the structural performance of reduced-beam section
(RBS) connections under alternated axial loading and lateral displacement, four full-scale specimens were tested. These tests were intended to assess the performance of the moment connection design for the Moscone Center Exp- ansion under the Design Basis Earthquake (DBE) and the Maximum Considered Earthquake (MCE). Previous research conducted on RBS moment connections [1,2] showed that connections with RBS profiles can achieve rotations in excess of 0.03 rad. However, doubts have been cast on the quality of the seismic performance of these connections under combined axial and lateral loading.
The Moscone Center Expansion is a three-story, 71,814 m2 (773,000 ft2) structure with steel moment frames as its primary lateral force-resisting system. A three dimensional perspective illustration is shown in Fig. 1. The overall height of the building, at the highest point of the exhibition roof, is approxima- tely 35.36 m (116ft) above ground level. The ceiling height at the exhibition hall is 8.23 m (27 ft) , and the typical floor-to-floor height in the building is 11.43 m (37.5 ft). The building was designed as type I according to the requi- rements of the 1997 Uniform Building Code. The framing system consists of four moment frames in the East–West direct- ion, one on either side of the stair towers, and four frames in the North–South direction, one on either side of the stair and elevator cores in the east end and two at the west end of the structure [4]. Because of the story height, the con- cept of the Coupled-Girder Moment-Resisting Framing System (CGMRFS) was utilized.
By coupling the girders, the lateral load-resisting behavior of the moment framing system changes to one where structural overturning moments are resisted partially by an axial compression–tension couple across the girder system, rather than only by the individual flexural action of the girders. As a result, a stiffer lateral load resisting system is achieved. The vertical element that connects the girders is referred to as a coupling link. Coupling links are analogous to and serve the same structural role as link beams in eccentrically braced frames. Coupling links are generally quite short, having a large shear- to-moment ratio.
Under earthquake-type loading, the CGMRFS subjects its girders to wariab- ble axial forces in addition to their end moments. The axial forces in the
Fig. 1. Moscone Center Expansion Project in San Francisco, CA
girders result from the accumulated shear in the link.
2. Analytical model of CGMRF
Nonlinear static pushover analysis was conducted on a typical one-bay model of the CGMRF. Fig. 2 shows the dimensions and the various sections of the 10 in) and the ?254 mm (1 1/8 in ?model. The link flange plates were 28.5 mm 18 3/4 in). The SAP 2000 computer ?476 mm (3 /8 in ?web plate was 9.5 mm program was utilized in the pushover analysis [5]. The frame was characterized as fully restrained(FR). FR moment frames are those frames for 1170 which no more than 5% of the lateral deflections arise from connection deformation [6]. The 5% value refers only to deflection due to beam–column deformation and not to frame deflections that result from column panel zone deformation [6, 9].
The analysis was performed using an expected value of the yield stress and ultimate strength. These values were equal to 372 MPa (54 ksi) and 518 MPa (75 ksi), respective ly. The plastic hinges’ load–deformation behavior was approximated by the generalized curve suggested by NEHRP Guidelines for the Seismic Rehabilitation of Buildings [6] as shown in Fig. 3. △y was calcu- lated based on Eqs. (5.1) and (5.2) from [6], as follows:
P–M hinge load–deformation model points C, D and E are based on Table 5.4 from [6] for
△y was taken as 0.01 rad per Note 3 in [6], Table 5.8. Shear hinge load- load–deformation model points C, D and E are based on Table 5.8 [6], Link Beam, Item a. A strain hardening slope between points B and C of 3% of the elastic slope was assumed for both models.
The following relationship was used to account for moment–axial load interaction [6]:
where MCE is the expected moment strength, ZRBS is the RBS plastic section modulus (in3), is the expected yield strength of the material (ksi), P is the axial force in the girder (kips) and is the expected axial yield force of the RBS, equal to (kips). The ultimate flexural capacities of the beam and the link of the model are shown in Table 1.
Fig. 4 shows qualitatively the distribution of the bending moment, shear force, and axial force in the CGMRF under lateral load. The shear and axial force in the beams are less significant to the response of the beams as compared with the bending moment, although they must be considered in design. The qualita- tive distribution of internal forces illustrated in Fig. 5 is fundamentally the same for both elastic and inelastic ranges of behavior. The specific values of the internal forces will change as elements of the frame yield and internal for- ces are redistributed. The basic patterns illustrated in Fig. 5, however, remain the same.
Inelastic static pushover analysis was carried out by applying monotonically increasing lateral displacements, at the top of both columns, as shown in Fig. 6. After the four RBS have yielded simultaneously, a uniform yielding in the web and at the
ends of the flanges of the vertical link will form. This is the yield mechanism for the frame , with plastic hinges also forming at the base of the columns if they are fixed. The base shear versus drift angle of the model is shown in Fig. 7 . The sequence of inelastic activity in the frame is shown on the figure. An elastic component, a long transition (consequence of the beam plastic hinges being formed simultaneously) and a narrow yield plateau characterize the pushover curve.
The plastic rotation capacity, qp, is defined as the total plastic rotation beyond which the connection strength starts to degrade below 80% [7]. This definition is different from that outlined in Section 9 (Appendix S) of the AISC Seismic Provisions [8, 10]. Using Eq. (2) derived by Uang and Fan [7], an estimate of the RBS plastic rotation capacity was found to be 0.037 rad:
Fyf was substituted for Ry?Fy [8], where Ry is used to account for the differ- ence between the nominal and the expected yield strengths (Grade 50 steel, Fy=345 MPa and Ry =1.1 are used).
3. Experimental program
The experimental set-up for studying the behavior of a connection was based on Fig. 6(a). Using the plastic displacement dp, plastic rotation gp, and plastic story drift angle qp shown in the figure, from geometry, it follows that:
And:
in which d and g include the elastic components. Approximations as above are used for large inelastic beam deformations. The diagram in Fig. 6(a) suggest that a sub assemblage with displacements controlled in the manner shown in Fig. 6(b) can represent the inelastic behavior of a typical beam in a CGMRF.
The test set-up shown in Fig. 8 was constructed to develop the mechanism shown in Fig. 6(a) and (b). The axial actuators were attached to three 2438 mm × 1219 mm ×1219 mm (8 ft × 4 ft × 4 ft) RC blocks. These blocks were tensioned to the laboratory floor by means of twenty-four 32 mm diameter dywidag rods. This arrangement permitted replacement of the specimen after each test.
Therefore, the force applied by the axial actuator, P, can be resolved into two or thogonal components, Paxial and Plateral. Since the inclination angle of the axial actuator does not exceed , therefore Paxial is approximately equal to P [4]. However, the lateral 3.0 component, Plateral, causes an additional moment at the beam-to column joint. If the axial actuators compress the specimen, then the lateral components will be adding to the lateral actuator forces, while if the axial actuators pull the specimen, the Plateral will be an opposing force to the lateral actuators. When the axial actuators undergo
axial actuators undergo a lateral displacement _, they cause an additional moment at the beam-to-column joint (P-△effect). Therefore, the moment at the beam-to column joint is equal to:
where H is the lateral forces, L is the arm, P is the axial force and _ is the lateral displacement.
Four full-scale experiments of beam column connections were conducted.
The member sizes and the results of tensile coupon tests are listed in Table 2
All of the columns and beams were of A572 Grade 50 steel (Fy 344.5 MPa). The actual measured beam flange yield stress value was equal to 372 MPa (54 ksi), while the ultimate strength ranged from 502 MPa (72.8 ksi) to 543 MPa (78.7 ksi).
Table 3 shows the values of the plastic moment for each specimen (based on measured tensile coupon data) at the full cross-section and at the reduced section at mid-length of the RBS cutout.
The specimens were designated as specimen 1 through specimen 4. Test specimens details are shown in Fig. 9 through Fig. 12. The following features were utilized in the design of the beam–column connection:
The use of RBS in beam flanges. A circular cutout was provided, as illustr- ated in Figs. 11 and 12. For all specimens, 30% of the beam flange width was removed. The cuts were made carefully, and then ground smooth in a direct- tion parallel to the beam flange to minimize notches.
Use of a fully welded web connection. The connection between the beam web and the column flange was made with a complete joint penetration groove weld (CJP). All CJP welds were performed according to AWS D1.1 Structural Welding Code
Use of two side plates welded with CJP to exterior sides of top and bottom beam flan- ges, from the face of the column flange to the beginning of the RBS, as shown in Figs. 11 and 12. The end of the side plate was smoothed to meet the beginning of the RBS. The side plates were welded with CJP with the column flanges. The side plate was added to increase the flexural capacity at the joint location, while the smooth transition was to reduce the stress raisers, which may initiate fracture
Two longitudinal stiffeners, 95 mm × 35 mm (3 3/4 in × 1 3/8 in), were welded with 12.7 mm (1/2 in) fillet weld at the middle height of the web as shown in Figs. 9 and 10. The stiffeners were welded with CJP to column flanges.
Removal of weld tabs at both the top and bottom beam flange groove welds. The weld tabs were removed to eliminate any potential notches introduced by the tabs or by weld discontinuities in the groove weld run out regions
Use of continuity plates with a thickness approximately equal to the beam flange thickness. One-inch thick continuity plates were used for all specimens.
While the RBS is the most distinguishing feature of these test specimens, the longitudinal stiffener played an important role in delaying the formation of web local buckling and developing reliable connection
4. Loading history
Specimens were tested by applying cycles of alternated load with tip displacement increments of _y as shown in Table 4. The tip displacement of the beam was imposed by servo-controlled actuators 3 and 4. When the axial force was to be applied, actuators 1 and 2 were activated such that its force simulates the shear force in the link to be transferred to the beam. 0.5_y. After The variable axial force was increased up to 2800 kN (630 kip) at that, this lo- ad was maintained constant through the maximum lateral displacement.
maximum lateral displacement. As the specimen was pushed back the axial
force remained constant until 0.5 y and then started to decrease to zero as the specimen passed through the neutral position [4]. According to the upper bound for beam axial force as discussed in Section 2 of this paper, it was concluded that P =2800 kN (630 kip) is appropriate to investigate this case in RBS loading. The tests were continued until failure of the specimen, or until limitations of the test set-up were reached.
5. Test results
The hysteretic response of each specimen is shown in Fig. 13 and Fig. 16. These plots show beam moment versus plastic rotation. The beam moment is measured at the middle of the RBS, and was computed by taking an equiva- lent beam-tip force multiplied by the distance between the centerline of the lateral actuator to the middle of the RBS (1792 mm for specimens 1 and 2, 3972 mm for specimens 3 and 4). The equivalent lateral force accounts for the additional moment due to P–△effect. The rotation angle was defined as the lateral displacement of the actuator divided by the length between the centerline of the lateral actuator to the mid length of the RBS. The plastic rotation was computed as follows [4]:
where V is the shear force, Ke is the ratio of V/q in the elastic range. Measurements and observations made during the tests indicated that all of the plastic rotation in specimen 1 to specimen 4 was developed within the beam. The connection panel zone and the column remained elastic as intended by design.
5.1. Specimens 1 and 2
The responses of specimens 1 and 2 are shown in Fig. 13. Initial yielding occurred during cycles 7 and 8 at 1_y with yielding observed in the bottom flange. For all test specimens, initial yielding was observed at this location and attributed to the moment at the base of the specimen [4]. Progressing through the loading history, yielding started to propagate along the RBS bottom flange. During cycle 3.5_y initiation of web buckling was noted adjacent to the yielded bottom flange. Yielding started to propagate along the top flange of the RBS and some minor yielding along the middle stiffener. During the cycle of 5_y with the increased axial compression load to 3115 KN (700 kips) a severe web buckle developed along with flange local buckling. The flange and the web local buckling became more pronounced with each successive loading cycle. It should be noted here that the bottom flange and web local buckling was not accompanied by a significant deterioration in the hysteresis loops.
A crack developed in specimen 1 bottom flange at the end of the RBS where it meets the side plate during the cycle 5.75_y. Upon progressing through the loading history, 7_y, the crack spread rapidly across the entire width of the bottom flange. Once the bottom flange was completely fractured, the web began to fracture. This fracture appeared to initiate at the end of the RBS，then propagated through the web net section of the shear tab, through the middle stiffener and the through the web net section on the other side of the stiffener. The maximum bending moment achieved on specimen 1 during the
During the cycle 6.5 y, specimen 2 also showed a crack in the bottom flange at the end of the RBS where it meets the wing plate. Upon progressing thou- gh the loading history, 15 y, the crack spread slowly across the bottom flan- ge. Specimen 2 test was stopped at this point because the limitation of the test set-up was reached.
The maximum force applied to specimens 1 and 2 was 890 kN (200 kip). The kink that is seen in the positive quadrant is due to the application of the varying axial tension force. The load-carrying capacity in this zone did not deteriorate as evidenced with the positive slope of the force–displacement curve. However, the load-carrying capacity deteriorated slightly in the neg- ative zone due to the web and the flange local buckling.
Photographs of specimen 1 during the test are shown in Figs. 14 and 15. Severe local buckling occurred in the bottom flange and portion of the web next to the bottom flange as shown in Fig. 14. The length of this buckle extended over the entire length of the RBS. Plastic hinges developed in the RBS with extensive yielding occurring in the beam flanges as well as the web. Fig. 15 shows the crack that initiated along the transition of the RBS to the side wing plate. Ultimate fracture of specimen 1 was caused by a fracture in the bottom flange. This fracture resulted in almost total loss of the beam- carrying capacity. Specimen 1 developed 0.05 rad of plastic rotation and showed no sign of distress at the face of the column as shown in Fig. 15.
5.2. Specimens 3 and 4
The response of specimens 3 and 4 is shown in Fig. 16. Initial yielding occured during cycles 7 and 8 at 1_y with significant yielding observed in the bottom flange. Progressing through the loading history, yielding started to propagate along the bottom flange on the RBS. During cycle 1.5_y initiation of web buckling was noted adjacent to the yielded bottom flange. Yielding started to propagate along the top flange of the RBS and some minor yielding along the middle stiffener. During the cycle of 3.5_y a severe web buckle developed along with flange local buckling. The flange and the web local buckling bec- ame more pronounced with each successive loading cycle.
During the cycle 4.5 y, the axial load was increased to 3115 KN (700 kips) causing yielding to propagate to middle transverse stiffener. Progressing through the loading history, the flange and the web local buckling became more severe. For both specimens, testing was stopped at this point due to limitations in the test set-up. No failures occurred in specimens 3 and 4. However, upon removing specimen 3 to outside the laboratory a hairline crack was observed at the CJP weld of the bottom flange to the column.
The maximum forces applied to specimens 3 and 4 were 890 kN (200 kip) and 912 kN (205 kip). The load-carrying capacity deteriorated by 20% at the end of the tests for negative cycles due to the web and the flange local buckling. This gradual reduction started after about 0.015 to 0.02 rad of plastic rotation. The load-carrying capacity during positive cycles (axial tension applied in the girder) did not deteriorate as evidenced with the slope of the force–displacement envelope for specimen 3 shown in Fig. 17.
A photograph of specimen 3 before testing is shown in Fig. 18. Fig. 19 is a
Fig. 16. Hysteretic behavior of specimens 3 and 4 in terms of moment at middle RBS versus beam plastic rotation.
photograph of specimen 4 taken after the application of 0.014 rad displacem- ent cycles, showing yielding and local buckling at the hinge region. The beam web yielded over its full depth. The most intense yielding was observed in the web bottom portion, between the bottom flange and the middle stiffener. The web top portion also showed yielding, although less severe than within the bottom portion. Yielding was observed in the longitudinal stiffener. No yiel- ding was observed in the web of the column in the joint panel zone. The un- reduced portion of the beam flanges near the face of the column did not show yielding either. The maximum displacement applied was 174 mm, and the maximum moment at the middle of the RBS was 1.51 times the plastic mom ent capacity of the beam. The plastic hinge rotation reached was about 0.032 rad (the hinge is located at a distance 0.54d from the column surface,where d is the depth of the beam).
5.2.1. Strain distribution around connection
The strain distribution across the flanges–outer surface of specimen 3 is shown in Figs. 20 and 21. The readings and the distributions of the strains in specimens 1, 2 and 4 (not presented) showed a similar trend. Also the seque- nce of yielding in these specimens is similar to specimen 3.
The strain at 51 mm from the column in the top flange–outer surface remained below 0.2% during negative cycles. The top flange, at the same location, yielded in compression only The longitudinal strains along the centerline of the bottom–flange outer face are shown in Figs. 22 and 23 for positive and negative cycles, respectively. From Fig.23, it is found that the strain on the RBS becomes several times larg- er than that near the column after cycles at –1.5_y; this is responsible for the
flange local buckling. Bottom flange local buckling occurred when the average strain in the plate reached the strain-hardening value (esh _ 0.018) and the reduced-beam portion of the plate was fully yielded under longitudinal stresses and permitted the development of a full buckled wave.
5.2.2. Cumulative energy dissipated
The cumulative energy dissipated by the specimens is shown in Fig. 24. The cumulative energy dissipated was calculated as the sum of the areas enclosed the lateral load–lateral displacement hysteresis loops. Energy dissipation sta- rted to increase after cycle 12 at 2.5 y (Fig. 19). At large drift levels, energy dissipation augments significantly with small changes in drift. Specimen 2 dissipated more energy than specimen 1, which fractured at RBS transition. However, for both specimens the trend is similar up to cycles at q =0.04 rad
In general, the dissipated energy during negative cycles was 1.55 times bigger than that for positive cycles in specimens 1 and 2. For specimens 3 and 4 the dissipated energy during negative cycles was 120%, on the average, that of the positive cycles. The combined phenomena of yielding, strain hardening, in-plane and out- of-plane deformations, and local distortion all occurred soon after the bottom flange RBS yielded.
6. Conclusions
Based on the observations made during the tests, and on the analysis of the instrumentation, the following conclusions were developed:
1. The plastic rotation exceeded the 3% radians in all test specimens.
2. Plastification of RBS developed in a stable manner.
3. The overstrength ratios for the flexural strength of the test specimens were equal to 1.56 for specimen 1 and 1.51 for specimen
4. The flexural strength capacity was based on the nominal yield strength and on the FEMA-273 beam–column equation.
4. The plastic local buckling of the bottom flange and the web was not accompanied by a significant deterioration in the load-carrying capacity.
5. Although flange local buckling did not cause an immediate degradation of strength, it did induce web local buckling.
6. The longitudinal stiffener added in the middle of the beam web assisted in transferring the axial forces and in delaying the formation of web local buckling. How ever, this has caused a much higher overstrength ratio, which had a significant impact on the capacity design of the welded joints, panel zone and the column.
7. A gradual strength reduction occurred after 0.015 to 0.02 rad of plastic rotation during negative cycles. No strength degradation was observed during positive cycles.
8. Compression axial load under 0.0325Py does not affect substantially the connection deformation capacity.
9. CGMRFS with properly designed and detailed RBS connections is a reliable system to resist earthquakes.
Acknowledgements
Structural Design Engineers, Inc. of San Francisco financially supported the experimental program. The tests were performed in the Large Scale Structures Laboratory of the University of Nevada, Reno. The participation of Elizabeth Ware, Adrianne Dietrich and of the technical staff is gratefully acknowledged.
References
受弯钢框架结点在变化轴向荷载和侧向位移的作用下的周期性行为
摘要
这篇论文讨论的是在变化的轴向荷载和侧向位移的作用下，接受测试的四种受弯钢结点的周期性行为。梁的试样由变截面梁，翼缘以及纵向的加劲肋组成。受测试样加载轴向荷载和侧向位移用以模拟侧向荷载对组合梁抗弯系统的影响。实验结果表明试样在旋转角度超过0.03弧度后经历了从塑性到延性的变化。纵向加劲肋的存在帮助传递轴向荷载以及延缓腹板的局部弯曲。
1、引言
为了评价变截面梁（RBS）结点在轴向荷载和侧向位移下的结构性能，对四个全尺寸的样品进行了测试。这些测试打算评价为旧金山展览中心扩建设计的受弯结点在满足设计基本地震等级（DBE）和最大可能地震等级（MCE）下的性能。基于上述而做的对RBS受弯结点的研究指出RBS形式的结点能够获得超过0.03弧度的旋转角度。然而，有人对于这些结点在轴向和侧向荷载作用下的抗震性能质量提出了怀疑。
旧金山展览中心扩建工程是一个3层构造，并以钢受弯框架作为基本的侧向力抵抗系统。Fig.1是一幅三维透视图。建筑的总标高为展览厅屋顶的最高点，大致是35.36m（116ft）。展览厅天花板的高度是8.23m（27ft），层高为11.43m（37.5ft）。建筑物按照1997统一建筑规范设计。
框架系统由以下几部分组成：四个东西走向的受弯框架，每个电梯塔边各一个；四个南北走向的受弯框架，在每个楼梯和电梯井各一个的；整体分布在建筑物的东西两侧。考虑到层高的影响，提出了双梁抗弯框架系统的观念。
通过连接大梁，受弯框架系统的抵抗荷载的行为转化为结构倾覆力矩部分地被梁系统的轴向压缩-拉伸分担，而不是仅仅通过梁的弯曲。结果，达到了一个刚性侧向荷载抵抗系统。竖向部分与梁以联结杆的形式连接。联结杆在结构中模拟偏心刚性构架并起到与其相同的作用。通常地联结杆都很短，并有很大的剪弯比。在地震类荷载的作用下，CGMRFS梁的最终弯矩将考虑到可变轴向力的影响。梁中的轴向力是切向力连续积累的结果。
2．CGMRF的解析模型
非线性静力推出器模型是以典型的单间CGMRF模板为指导。图2展示了模型的尺寸规格和多个部分。翼缘板尺寸为28.5mm 254mm（1 1/8in 10in），腹板尺寸为9.5mm 476mm（3/8in 18 3/4in）。推进器模型中运用了SAP 2000计算机程序。框架的特色是全约束（FR）。FR受弯框架是一种由结点应变引起的挠度不超过侧向挠度的5%的框架。这个5%仅与梁-柱应变有关，而与柱底板区应变引起的框架应变无关。
模型通过屈服应力和匹配强度的期望值来运行。这些值各自为372Mpa（54ksi）和518Mpa（75ksi）。Fig.3显示了塑性铰的荷载-应变行为是通过建筑物地震恢复的NEHRP指标以广义曲线的形式逼近的。y以Eps5.1和5.2为基底运算，如下：
P-M铰合线荷载-应变模型上的点C，D和E的取值如表5.4
y以0.01rad为幅度取值见表5.8。切变铰合线荷载-应变模型点C，D和E取值见表5.8。对于连续梁，假定两个模型点B和C之间的形变硬化比有3%的弹性
比。
用下面的公式计算弯矩与轴向荷载之间的相互关系
是期望弯矩强度，是RBS塑性模量，是材料的屈服强度，P是梁中的轴向力，是RBS屈服力，等于。梁的最终弯曲能力和模型的连续行见图1。
Fig.4定性的给出了侧向荷载下的CGMRF中的弯矩，切应力和正应力的分布。其中切应力和正应力对梁的影响要小于弯矩的作用，尽管他们必须在设计中加以考虑。内力分布图解见Fig.5，可见，弹性范围和非弹性范围的内力行为基本相同。内力的比值将随框架的屈服和内力的重分布的变化而变化。基本内力图见Fig.5，然而，仍然是一样的。
非静力推进器模型的运行通过柱子顶部的侧向位移的单调增加来实现，如Fig.5所示。在四个RBS同时屈服后，发生在腹板与翼缘端部的竖向的统一屈服将开始形成。这是框架的屈服中心，在柱子被固定后将在柱底部形成塑性铰。Fig.7给出了基本切应力偏移角。图中还给出了框架中非弹性活动的次序。对于一个弹性组成，推进器将有一个特有的很长的过渡（同时形成塑性铰）和一个很短的屈服平稳阶段。
塑性旋转能力，被定义为：结点强度从开始递减到低于80%的总的塑性旋转角。这个定义不同于第9段（附录）AISC地震条款的描述。使用Eq源于RBS塑性旋转能力被定在0.037弧度。
被替代，用来计算理论屈服强度与实际屈服强度的区别（标号是50钢）Table 1
梁的弯曲能力和模型的连接：
3．实践规划
如图6所示，实验布置是为了研究基于典型的CGMRF结构下的结点在动力学中的能量耗散。用图中所给的塑性位移，塑性转角，塑性偏移角，由几何结构，有如下：和
这里的δ和γ包括了弹性组合。上述近似值用于大型的非弹性梁的变形破坏。图6a表明用图6b所示的位移控制下的替代组合能够表示CGMRF结构中的典型梁的非弹性行为。
图8所示，建立这个实验装置来发展图6a和图6b所示的机构学。轴心装置附以3个2438mm×1219mm×1219mm（8ft×4ft×4ft）RC块。并用24个32mm 径的杆与实验室的地板固定。这种装置允许在每次测验后换实验样品。
根据实验布置的动力学要求，随着侧面的元件放置，轴向的元件，元件1和元件2，将钉到B和C中去，如图8所示。因此，轴向元件提供的轴向力P可以被分解为相互正交的力的组合，和，由于轴向力的倾斜角度不超过，因此近似等于P。然而，侧向力分量，，引起了一个在梁柱交接处的附加弯矩。如果轴向元件压试样的话，那么将会加到侧向力中，若轴向是拉力，对于侧向元件来说则是个反向力。当轴向元件有个侧向位移，他们将在梁柱交接处引起一个附加弯矩，因此，梁柱交接处的弯矩等于：
M=HL+P
其中H是侧向力，L是力臂，P是轴向力，是侧向位移。
四个梁柱结点全尺寸实验做完了。拉伸试样检测的结果和构件尺寸见表2。所有柱和梁的钢筋为A572标号50钢（=344.5Mpa）。经测定的梁翼缘屈服应力值等于372Mpa（54ksi），整体的强度范围是从502Mpa（72.8ksi）到543Mpa （78.7ksi）。
表3列出了各个试样的全截面和RBS中间变截面处的塑性弯矩值（受拉应力下的数据）。
本文所指的试样专指试样1到4。被检试样细部图见图9到图12。在设计梁柱结点时用到了以下数据：
梁翼缘部分采用RBS结构。配备环形掏槽，如图11和图12所示。对于所有的试样，切除30%翼缘宽度。切除工作做的十分精细，并打磨光滑且与梁翼缘保持平行以尽量见效切口。
应用全焊接腹板结点。梁腹板与柱翼缘之间的结点采用全焊缝焊接（CJP）。所有CJP焊接严格依照AWS D1.1结构焊接规范。
采用双侧板加CJP形式连接梁翼缘的顶部和底部和柱表面到变截面开始处，如图11和图12。侧板尾部打磨光滑以便同RBS连接。侧板采用CJP形式与柱边缘相连接。侧板的作用是增加受弯单元的承受能力，平稳过渡是为了减少应力集中而导致的破裂。
两根纵向的加劲肋，95mm×35mm（3 3/4in ×1 3/8 in），以12.7mm的角焊缝焊接到腹板的中间高度，如图9和10。加劲肋采用CJP的形式焊接到柱的边缘。切除梁翼缘顶部和低部的坡口焊缝处的多余焊接部分。以便消除坡口焊接断口处可能产生的断裂。
除去翼缘低部的衬垫板条。以便消除衬垫板条带来的断口效应并增加安全性。使用与梁翼缘厚度近似相同的连续板。所有试样板厚均为一英寸。
由于RBS是受检试样最容易区分的特征，纵向的加劲肋在延缓局部弯曲和提高结点可靠性方面扮演着重要的角色。
4．荷载历史
试样被加以周期性交替的荷载，其末端的位移△y的增加如图4所示。梁的末端位移受伺服控制装置3和4的影响。当作用轴向力时，制动器1和2是活动的，以此用它的受力来模拟从连接处传到梁上的剪力。可变的轴向荷载在+0.5△y处增加到2800KN。在那以后，通过最大的侧向位移，这个荷载保持恒定。在试样被推回时，轴向力维持恒定直至0.5△y，然后减小到零，此时的试样通过中和轴。根据本文第2部分有关轴向力受以上约束的论述，可以推断出以P=2800KN 来研究RBS负载是合理的。测试将会继续，直至试样损坏，或者到实验预期的限制。
5.实验结果
每个试样的滞后反应见图13和图16。这些图表显示了梁弯矩相对的的塑性旋度。梁的弯矩在RBS试样的中间测量，并通过取一个等价的梁端力乘以制动器侧向中心线到RBS中间的距离来计算。（试样1和2为1792mm，试样3和4为23972mm）。用来计算附加弯矩的等价侧向力由于P-△。旋转角是这样定义的，用制动器的侧向位移除以制动器侧向中心线到RBS中间的距离。塑性旋度计算如下：
其中V是剪力，K是弹性在范围内的比。在测试期间的测量和观察表明，试样1和4的所有塑性旋度均在梁的内部发展。板的连接区域和柱子保持弹性，如设计预期的一样。
表5列出了每个试样在测试最后的塑性旋度。塑性旋度合格性能的目标级被定在±0.03rad，依AISC钢结构建筑抗震条例而定。所有试样均达到了合格的性能标准。
所有试样均有良好的塑性变形和能量耗散。当负载周期为±1△y时，底部首先屈服，然后随着负载周期逐渐扩散增加。
5.1 试样1和2
试样1和2的变化见图13。在第7和第8个周期以及1△y，最初屈服发生在底缘处。对于所有的受测试的试样，最初的屈服均发生在这个部位，这是由试样底部的弯矩引起的。随着荷载作用的继续，屈服开始沿着RBS底缘传递。从3.5△y开始，发生腹板弯曲并且相邻的底缘开始屈服。屈服开始沿RBS上边缘传递，一些次要的屈服传递到中间的加劲肋。在5△y开始，轴向压力增大到3115KN，一个剧烈的腹板的翘曲产生并伴随着局部弯曲。腹板和翼缘的局部弯曲随着荷载的累次加载而逐渐明显。这里要说明的是，在滞后回线中，腹板和翼缘的局部弯曲并没附有重要的损坏。
当作用到5.75△y时，在RBS的尾部和衬板连接处，试样1的底缘产生一个裂缝。随着荷载周期的增加到7△y时，裂缝迅速扩大并穿过了整个底缘。一旦底缘完全断裂，腹板将开始断裂。这个断裂首先在RBS的末端出现，然后沿剪切槽的净截面传播，通过加劲肋的中间并通过另一边的加劲肋的净截面。在实验中，试样1的最大作用弯矩是梁的塑性承载力的1.56倍。
在作用到6.5△y时，试样2也在底缘处出现一个裂缝，是在RBS末端与翼板的交接处。随着荷载周期的增加，第15△y时，裂缝缓慢的发展穿过了底缘。试样2的测试到此结束，因为已经到了实验装置加载的极限。
加给试样1和试样2的最大荷载是890KN。从正的象限中看到的弯折是由于施加的变化的轴向拉力导致。力-位移曲线的正斜率证明了这个区域的负载容量并没有减弱。然而，由于腹板和翼缘的局部弯曲的影响，负的区域的负载容量有轻微的削弱。
试样1的照片如图14和图15。由图14可以看到，底缘处发生严重的局部弯曲，并可以看到与底缘相连的腹板部分。弯曲沿展到整个RBS的长度方向。RBS中形成塑性胶，并伴随着梁的腹板和翼缘的大规模的屈服。由图15可见，裂缝由RBS的连接传递到了侧面的翼板。在底缘的一个断裂导致了试样1的最终断裂。这个断裂导致梁几乎失去承载能力。图15还说明了试样1产成了0.05rad的塑性旋度，并且在柱子表面没有疲劳损伤。
5.2 试样3和试样4
试样3和试样4的变化曲线如图16。最初的屈服发生在荷载周期第7到第8周之间，底缘的重要屈服发生在1△y处。随着荷载周期的发展，屈服开始沿RBS 的底缘传播。在1.5△y时，腹板弯曲发生并明显伴随着底缘的屈服。屈服开始沿着RBS的顶部传播，一些次要的屈服沿着加劲肋中部传播。在荷载周期到3.5△y时，一个剧烈的腹板翘曲产生并伴随着翼缘的局部弯曲。腹板和翼缘的局部弯曲随着累次加载变得逐渐明显。
当加载到4.5△y时，轴向荷载增大到3115KN，并导致屈服传播到中间横向加强构件。随着荷载周期的增加，腹板和翼缘的局部弯曲变得更加剧烈。对于2
个试样，受实验装置的约束测试到此结束。在试样3和试样4中没有破坏产生。然而，在将试样3移动到实验室之外时，却发现在底缘与柱子的焊接处有一个微小的裂缝。
加给试样3和试样4的最大荷载分别是890KN和912KN。试样的负载容量在实验后削弱了20%，这是由腹板和翼缘的局部弯曲引起的。这个慢性的恢复在大概塑性旋度产生0.015到0.02后开始。如图17所示，试样3在正的象限中的
负载容量没有减弱（轴向的拉伸作用在梁上），由力-位移的包络图可见。
图18是试样3的测试前的照片。图19是试样4在0.014的位移作用周期后的照片，显示了铰合区域的屈服和局部弯曲。梁的腹板的屈服沿着其整个深度方向。最强的屈服发生在腹板的底部，底缘和中间加劲肋之间。腹板的顶部也发生了屈服，虽然其剧烈程度不如底部。纵向的加劲肋也发生了屈服。柱子的连接板部分没有发生屈服。在接近柱子表面的梁的未经削弱的部分也没有显示发生屈服。最大位移是174mm，最大弯矩发生在RBS中部，为梁的塑性弯矩值的1.51倍。塑性铰的旋度达到了0.032rad（铰接点设置在距离柱子表面0.54d处，其中d 是梁的长度）。
5.2.1 结点处的应变分布
试样3的外表面边缘的应变分布见图20和21。试样1、2、4的应变记录和分布状态呈现了相似的趋势。同样的，这些试样的屈服次序也同试样3的相似。在负周期时，离柱子51mm的顶部外表面处的应变低于0.2%。位于顶部同一位置的翼缘，仅在受压时屈服。
图22和图23显示了沿着底缘外表面中心线的纵向应变，其中取22是正向周期下的，图23是负向周期下的。从图23我们可以看出，在周期在-1.5△y以后，RBS上的应变比附近的柱子上的应变要大好几倍；这是由翼缘的局部弯曲造成的。底缘局部弯曲发生在整个板的平均应变达到形变硬化值时，板的变截面部分在纵向力下完全屈服，从而导致一个十分弯曲的波纹。
5.2.2 累计能量的消散
试样的累计能量消散见图24。累计能量消散是以封入区域的横向荷载的滞后回线之和表示的。能量消散在加载到12周以后在2.5△y处开始增加。对于飘移电平，点平的很小变化会带来很大的能量耗散。试样2比试样1消耗更多的能量，它是在RBS过度部分断裂的。然而，对于2个试样来说，在θ=0.04rad时，其周期是相似的。
总的来说，在试样1和试样2中，负的周期下的能量消散比正的周期下能量消散大1.55倍。对于试样3和试样4来说，负的能量消散是正负平均水平的120%。在底缘RBS屈服后，屈服的组合现象，应变硬化，面内形变和面外形变，局部弯曲均很快发生。
6．结论
基于由实验而得的数据，以及应用于仪器的解析法，得出如下结论：
1、对于所有的试样，塑性旋度均超出0.3%。
2、 RBS的塑性过程是平稳发展的。
3、试样超出抗弯强度的比率，试样1等于1.56，试样4等于1.51。抗弯强度承载能力取决于标定的屈服强度和FEMA-273梁-柱等式。
4、底缘和腹板的塑性局部弯曲对其荷载承受能力没有重大的削弱。
5、尽管翼缘的局部弯曲不使强度立即产生削弱，但是它确实导致腹板的局部弯曲。
6、设置在梁的腹板中部的纵向加劲肋，能够帮助传递轴向力，还能延缓腹板的局部弯曲。然而，它却产生如此大的一个超过强度的比率，从而使焊接结点、板条区域以及柱子的承载能力在设计时大打折扣。
7、在负载周期时，塑性旋度为0.015到0.02rad时将会产生一个逐渐的强度的减弱。在正向周期时也没有。
8、轴想压缩荷载在小于0.0325Py时，对结点应变能力影响不大。
9、 CGMRFS技术与适当的设计以及详细的RBS连接，是一个可靠的抗震系统
PA VEMENT PROBLEMS CAUSED BY COLLAPSIBLE SUBGRADES By Sandra L. Houston,1 Associate Member, ASCE (Reviewed by the Highway Division) ABSTRACT: Problem subgrade materials consisting of collapsible soils are com- mon in arid environments, which have climatic conditions and depositional and weathering processes favorable to their formation. Included herein is a discussion of predictive techniques that use commonly available laboratory equipment and testing methods for obtaining reliable estimates of the volume change for these problem soils. A method for predicting relevant stresses and corresponding collapse strains for typical pavement subgrades is presented. Relatively simple methods of evaluating potential volume change, based on results of familiar laboratory tests, are used. INTRODUCTION When a soil is given free access to water, it may decrease in volume, increase in volume, or do nothing. A soil that increases in volume is called a swelling or expansive soil, and a soil that decreases in volume is called a collapsible soil. The amount of volume change that occurs depends on the soil type and structure, the initial soil density, the imposed stress state, and the degree and extent of wetting. Subgrade materials comprised of soils that change volume upon wetting have caused distress to highways since the be- ginning of the professional practice and have cost many millions of dollars in roadway repairs. The prediction of the volume changes that may occur in the field is the first step in making an economic decision for dealing with these problem subgrade materials. Each project will have different design considerations, economic con- straints, and risk factors that will have to be taken into account. However, with a reliable method for making volume change predictions, the best design relative to the subgrade soils becomes a matter of economic comparison, and a much more rational design approach may be made. For example, typical techniques for dealing with expansive clays include: (1) In situ treatments with substances such as lime, cement, or fly-ash; (2) seepage barriers and/ or drainage systems; or (3) a computing of the serviceability loss and a mod- ification of the design to "accept" the anticipated expansion. In order to make the most economical decision, the amount of volume change (especially non- uniform volume change) must be accurately estimated, and the degree of road roughness evaluated from these data. Similarly, alternative design techniques are available for any roadway problem. The emphasis here will be placed on presenting economical and simple methods for: (1) Determining whether the subgrade materials are collapsible; and (2) estimating the amount of volume change that is likely to occur in the 'Asst. Prof., Ctr. for Advanced Res. in Transp., Arizona State Univ., Tempe, AZ 85287. Note. Discussion open until April 1, 1989. To extend the closing date one month,
本科毕业设计 外文文献及译文 文献、资料题目：Designing Against Fire Of Building 文献、资料来源：国道数据库 文献、资料发表（出版）日期：2008.3.25 院（部）：土木工程学院 专业：土木工程 班级：土木辅修091 姓名：武建伟 学号：2008121008 指导教师：周学军、李相云 翻译日期： 20012.6.1
外文文献: Designing Against Fire Of Buliding John Lynch ABSTRACT: This paper considers the design of buildings for fire safety. It is found that fire and the associ- ated effects on buildings is significantly different to other forms of loading such as gravity live loads, wind and earthquakes and their respective effects on the building structure. Fire events are derived from the human activities within buildings or from the malfunction of mechanical and electrical equipment provided within buildings to achieve a serviceable environment. It is therefore possible to directly influence the rate of fire starts within buildings by changing human behaviour, improved maintenance and improved design of mechanical and electrical systems. Furthermore, should a fire develops, it is possible to directly influence the resulting fire severity by the incorporation of fire safety systems such as sprinklers and to provide measures within the building to enable safer egress from the building. The ability to influence the rate of fire starts and the resulting fire severity is unique to the consideration of fire within buildings since other loads such as wind and earthquakes are directly a function of nature. The possible approaches for designing a building for fire safety are presented using an example of a multi-storey building constructed over a railway line. The design of both the transfer structure supporting the building over the railway and the levels above the transfer structure are considered in the context of current regulatory requirements. The principles and assumptions associ- ated with various approaches are discussed. 1 INTRODUCTION Other papers presented in this series consider the design of buildings for gravity loads, wind and earthquakes.The design of buildings against such load effects is to a large extent covered by engineering based standards referenced by the building regulations. This is not the case, to nearly the same extent, in the
转型衰退时期的土木工程研究 Sergios Lambropoulosa[1], John-Paris Pantouvakisb, Marina Marinellic 摘要 最近的全球经济和金融危机导致许多国家的经济陷入衰退，特别是在欧盟的周边。这些国家目前面临的民用建筑基础设施的公共投资和私人投资显著收缩，导致在民事特别是在民用建筑方向的失业。因此，在所有国家在经济衰退的专业发展对于土木工程应届毕业生来说是努力和资历的不相称的研究，因为他们很少有机会在实践中积累经验和知识，这些逐渐成为过时的经验和知识。在这种情况下，对于技术性大学在国家经济衰退的计划和实施的土木工程研究大纲的一个实质性的改革势在必行。目的是使毕业生拓宽他们的专业活动的范围，提高他们的就业能力。 在本文中，提出了土木工程研究课程的不断扩大，特别是在发展的光毕业生的潜在的项目，计划和投资组合管理。在这个方向上，一个全面的文献回顾，包括ASCE体为第二十一世纪，IPMA的能力的基础知识，建议在其他：显著增加所提供的模块和项目管理在战略管理中添加新的模块，领导行为，配送管理，组织和环境等；提供足够的专业训练五年的大学的研究；并由专业机构促进应届大学生认证。建议通过改革教学大纲为土木工程研究目前由国家技术提供了例证雅典大学。 1引言 土木工程研究（CES）蓬勃发展，是在第二次世界大战后。土木工程师的出现最初是由重建被摧毁的巨大需求所致，目的是更多和更好的社会追求。但是很快，这种演变一个长期的趋势，因为政府为了努力实现经济发展，采取了全世界的凯恩斯主义的理论，即公共基础设施投资作为动力。首先积极的结果导致公民为了更好的生活条件（住房，旅游等）和增加私人投资基础设施而创造机会。这些现象再国家的发展中尤为为明显。虽然前景并不明朗（例如，世界石油危机在70年代），在80年代领先的国家采用新自由主义经济的方法（如里根经济政策），这是最近的金融危机及金融危机造成的后果（即收缩的基础设施投资，在技术部门的高失业率），消除发展前途无限的误区。 技术教育的大学所认可的大量研究土木工程部。旧学校拓展专业并且新的学校建成，并招收许多学生。由于高的职业声望，薪酬，吸引高质量的学校的学生。在工程量的增加和科学技术的发展，导致到极强的专业性，无论是在研究还是工作当中。结构工程师，液压工程师，交通工程师等，都属于土木工程。试图在不同的国家采用专业性的权利，不同的解决方案，，从一个统一的大学学历和广泛的专业化的一般职业许可证。这个问题在许多其他行业成为关键。国际专业协会的专家和机构所确定的国家性检查机构，经过考试后，他们证明不仅是行业的新来者，而且专家通过时间来确定进展情况。尽管在很多情况下，这些证书虽然没有国家接受，他们赞赏和公认的世界。 在试图改革大学研究（不仅在土木工程）更接近市场需求的过程中，欧盟确定了1999博洛尼亚宣言，它引入了一个二能级系统。第一级度（例如，一个三年的学士）是进入
零售企业营销策略中英文对照外文翻译文献(文档含英文原文和中文翻译)
译文： 零售企业的营销策略 Philip Kotlor 今天的零售商为了招徕和挽留顾客，急欲寻找新的营销策略。过去，他们挽留顾客的方法是销售特别的或独特的花色品种，提供比竞争对手更多更好的服务提供商店信用卡是顾客能赊购商品。可是，现在这一切都已变得面目全非了。现在，诸如卡尔文·克连，依佐和李维等全国性品牌，不仅在大多数百货公司及其专营店可以看到，并且也可以在大型综合商场和折扣商店可以买到。全国性品牌的生产商为全力扩大销售量，它们将贴有品牌的商品到处销售。结果是零售商店的面貌越来越相似。 在服务项目上的分工差异在逐渐缩小。许多百货公司削减了服务项目，而许多折扣商店却增加了服务项目。顾客变成了精明的采购员，对价格更加敏感。他们看不出有什么道理要为相同的品牌付出更多的钱，特别是当服务的差别不大或微不足道时。由于银行信用卡越来越被所有的商家接受，他们觉得不必从每个商店赊购商品。 百货商店面对着日益增加的价格的折扣店和专业商店的竞争，准备东山再起。历史上居于市中心的许多商店在郊区购物中心开设分店，那里有宽敞的停车场，购买者来自人口增长较快并且有较高收入的地区。其他一些则对其商店形式进行改变，有些则试用邮购盒电话订货的方法。超级市场面对的是超级商店的竞争，它们开始扩大店面，经营大量的品种繁多的商品和提高设备等级，超级市场还增加了它们的促销预算，大量转向私人品牌，从而增加盈利。 现在，我们讨论零售商在目标市场、产品品种和采办、服务以及商店气氛、定价、促销和销售地点等方面的营销策略。 一、目标市场 零售商最重要的决策时确定目标市场。当确定目标市场并且勾勒出轮廓时，零售商才能对产品分配、商店装饰、广告词和广告媒体、价格水平等作出一致的决定。如沃尔玛的目标市场相当明确：
姓名： 学号： 10447425 X X 大学 毕业设计（论文）外文翻译 （2014届） 外文题目Developments in excavation bracing systems 译文题目开挖工程支撑体系的发展 外文出处Tunnelling and Underground Space Technology 31 (2012) 107–116 学生XXX 学院XXXX 专业班级XXXXX 校内指导教师XXX 专业技术职务XXXXX 校外指导老师专业技术职务 二○一三年十二月
开挖工程支撑体系的发展 1.引言 几乎所有土木工程建设项目（如建筑物，道路，隧道，桥梁，污水处理厂，管道，下水道）都涉及泥土挖掘的一些工程量。往往由于由相邻的结构，特性线，或使用权空间的限制，必须要一个土地固定系统，以允许土壤被挖掘到所需的深度。历史上，许多挖掘支撑系统已经开发出来。其中，现在比较常见的几种方法是：板桩，钻孔桩墙，泥浆墙。 土地固定系统的选择是由技术性能要求和施工可行性（例如手段，方法）决定的，包括执行的可靠性，而成本考虑了这些之后，其他问题也得到解决。通常环境后果（用于处理废泥浆和钻井液如监管要求）也非常被关注（邱阳、1998）。 土地固定系统通常是建设项目的较大的一个组成部分。如果不能按时完成项目，将极大地影响总成本。通常首先建造支撑，在许多情况下，临时支撑系统是用于支持在挖掘以允许进行不断施工，直到永久系统被构造。临时系统可以被去除或留在原处。 打桩时，因撞击或振动它们可能会被赶入到位。在一般情况下，振动是最昂贵的方法，但只适合于松散颗粒材料，土壤中具有较高电阻（例如，通过鹅卵石）的不能使用。采用打入桩系统通常是中间的成本和适合于软沉积物（包括粘性和非粘性），只要该矿床是免费的鹅卵石或更大的岩石。 通常，垂直元素（例如桩）的前安装挖掘工程和水平元件（如内部支撑或绑回）被安装为挖掘工程的进行下去，从而限制了跨距长度，以便减少在垂直开发弯矩元素。在填充情况下，桩可先设置，从在斜坡的底部其嵌入悬挑起来，安装作为填充进步水平元素（如搭背或土钉）。如果滞后是用来保持垂直元素之间的土壤中，它被安装为挖掘工程的进行下去，或之前以填补位置。 吉尔- 马丁等人（2010）提供了一个数值计算程序，以获取圆形桩承受轴向载荷和统一标志（如悬臂桩）的单轴弯矩的最佳纵筋。他们开发的两种优化流程：用一个或两个直径为纵向钢筋。优化增强模式允许大量减少的设计要求钢筋的用量，这些减少纵向钢筋可达到50％相对传统的，均匀分布的加固方案。 加固桩集中纵向钢筋最佳的位置在受拉区。除了节约钢筋，所述非对称加强钢筋图案提高抗弯刚度，通过增加转动惯量的转化部分的时刻。这种增加的刚性可能会在一段时间内增加的变形与蠕变相关的费用。评估相对于传统的非对称加强桩的优点，对称，钢筋桩被服务的条件下全面测试来完成的，这种试验是为了验证结构的可行性和取得的变形的原位测量。 基于现场试验中，用于优化的加强图案的优点浇铸钻出孔（CIDH）在巴塞罗那的
专业资料 学院： 专业：土木工程 姓名： 学号： 外文出处：Structural Systems to resist (用外文写) Lateral loads 附件：1.外文资料翻译译文；2.外文原文。
附件1：外文资料翻译译文 抗侧向荷载的结构体系 常用的结构体系 若已测出荷载量达数千万磅重，那么在高层建筑设计中就没有多少可以进行极其复杂的构思余地了。确实，较好的高层建筑普遍具有构思简单、表现明晰的特点。 这并不是说没有进行宏观构思的余地。实际上，正是因为有了这种宏观的构思，新奇的高层建筑体系才得以发展，可能更重要的是：几年以前才出现的一些新概念在今天的技术中已经变得平常了。 如果忽略一些与建筑材料密切相关的概念不谈，高层建筑里最为常用的结构体系便可分为如下几类： 1.抗弯矩框架。 2.支撑框架，包括偏心支撑框架。 3.剪力墙，包括钢板剪力墙。 4.筒中框架。 5.筒中筒结构。 6.核心交互结构。 7. 框格体系或束筒体系。 特别是由于最近趋向于更复杂的建筑形式，同时也需要增加刚度以抵抗几力和地震力，大多数高层建筑都具有由框架、支撑构架、剪力墙和相关体系相结合而构成的体系。而且，就较高的建筑物而言，大多数都是由交互式构件组成三维陈列。 将这些构件结合起来的方法正是高层建筑设计方法的本质。其结合方式需要在考虑环境、功能和费用后再发展，以便提供促使建筑发展达到新高度的有效结构。这并
不是说富于想象力的结构设计就能够创造出伟大建筑。正相反，有许多例优美的建筑仅得到结构工程师适当的支持就被创造出来了，然而，如果没有天赋甚厚的建筑师的创造力的指导，那么，得以发展的就只能是好的结构，并非是伟大的建筑。无论如何，要想创造出高层建筑真正非凡的设计，两者都需要最好的。 虽然在文献中通常可以见到有关这七种体系的全面性讨论，但是在这里还值得进一步讨论。设计方法的本质贯穿于整个讨论。设计方法的本质贯穿于整个讨论中。 抗弯矩框架 抗弯矩框架也许是低，中高度的建筑中常用的体系，它具有线性水平构件和垂直构件在接头处基本刚接之特点。这种框架用作独立的体系，或者和其他体系结合起来使用，以便提供所需要水平荷载抵抗力。对于较高的高层建筑，可能会发现该本系不宜作为独立体系，这是因为在侧向力的作用下难以调动足够的刚度。 我们可以利用STRESS，STRUDL 或者其他大量合适的计算机程序进行结构分析。所谓的门架法分析或悬臂法分析在当今的技术中无一席之地，由于柱梁节点固有柔性，并且由于初步设计应该力求突出体系的弱点，所以在初析中使用框架的中心距尺寸设计是司空惯的。当然，在设计的后期阶段，实际地评价结点的变形很有必要。 支撑框架 支撑框架实际上刚度比抗弯矩框架强，在高层建筑中也得到更广泛的应用。这种体系以其结点处铰接或则接的线性水平构件、垂直构件和斜撑构件而具特色，它通常与其他体系共同用于较高的建筑，并且作为一种独立的体系用在低、中高度的建筑中。
项目成本控制 一、引言 项目是企业形象的窗口和效益的源泉。随着市场竞争日趋激烈，工程质量、文明施工要求不断提高，材料价格波动起伏，以及其他种种不确定因素的影响，使得项目运作处于较为严峻的环境之中。由此可见项目的成本控制是贯穿在工程建设自招投标阶段直到竣工验收的全过程，它是企业全面成本管理的重要环节，必须在组织和控制措施上给于高度的重视，以期达到提高企业经济效益的目的。 二、概述 工程施工项目成本控制，指在项目成本在成本发生和形成过程中，对生产经营所消耗的人力资源、物资资源和费用开支，进行指导、监督、调节和限制，及时预防、发现和纠正偏差从而把各项费用控制在计划成本的预定目标之内，以达到保证企业生产经营效益的目的。 三、施工企业成本控制原则 施工企业的成本控制是以施工项目成本控制为中心，施工项目成本控制原则是企业成本管理的基础和核心，施工企业项目经理部在对项目施工过程进行成本控制时，必须遵循以下基本原则。 3.1 成本最低化原则。施工项目成本控制的根本目的，在于通过成本管理的各种手段，促进不断降低施工项目成本，以达到可能实现最低的目标成本的要求。在实行成本最低化原则时，应注意降低成本的可能性和合理的成本最低化。一方面挖掘各种降低成本的能力，使可能性变为现实；另一方面要从实际出发，制定通过主观努力可能达到合理的最低成本水平。 3.2 全面成本控制原则。全面成本管理是全企业、全员和全过程的管理，亦称“三全”管理。项目成本的全员控制有一个系统的实质性内容，包括各部门、各单位的责任网络和班组经济核算等等，应防止成本控制人人有责，人人不管。项目成本的全过程控制要求成本控制工作要随着项目施工进展的各个阶段连续 进行，既不能疏漏，又不能时紧时松，应使施工项目成本自始至终置于有效的控制之下。 3.3 动态控制原则。施工项目是一次性的，成本控制应强调项目的中间控制，即动态控制。因为施工准备阶段的成本控制只是根据施工组织设计的具体内容确
顾客满意策略与顾客满意营销 原文来源：《Marketing Customer Satisfaction 》自20世纪八十年代末以来, 顾客满意战略已日益成为各国企业占有更多的顾客份额, 获得竞争优势的整体经营手段。 一、顾客满意策略是现代企业获得顾客“货币选票”的法宝随着时代的变迁, 社会物质财富的极大充裕, 顾客中的主体———消费者的需求也先后跨越了物质缺乏的时代、追求数量的时代、追求品质的时代, 到了20世纪八十年代末进入了情感消费时代。在我国, 随着经济的高速发展,我们也已迅速跨越了物质缺乏时代、追求数量的时代乃至追求品质的时代, 到今天也逐步迈进情感消费时代。在情感消费时代, 各企业的同类产品早已达到同时、同质、同能、同价, 消费者追求的已不再是质量、功能和价格, 而是舒适、便利、安全、安心、速度、跃动、环保、清洁、愉快、有趣等,消费者日益关注的是产品能否为自己的生活带来活力、充实、舒适、美感和精神文化品位, 以及超越消费者期望值的售前、售中、售后服务和咨询。也就是说, 今天人们所追求的是具有“心的满足感和充实感”的商品, 是高附加值的商品和服务,追求价值观和意识多元化、个性化和无形的满足感的时代已经来临。 与消费者价值追求变化相适应的企业间的竞争, 也由产品竞争、价格竞争、技术竞争、广告竞争、品牌竞争发展到现今的形象竞争、信誉竞争、文化竞争和服务竞争, 即顾客满意竞争。这种竞争是企业在广角度、宽领域的时空范围内展开的高层次、体现综合实力的竞争。它包括组织创新力、技术创新力、管理创新力、产业预见力、产品研发力、员工向心力、服务顾客力、顾客亲和力、同行认同力、社会贡献力、公关传播沟通力、企业文化推动力、环境适应力等等。这些综合形象力和如何合成综合持久的竞争力, 这就是CSft略所要解决的问题。CS寸代,企业不再以“自己为中心”,而是以“顾客为中心”；“顾客为尊”、“顾客满意”不再是流于形式的口号, 而是以实实在在的行动为基础的企业经营的一门新哲学。企业不再以质量达标, 自己满意为经营理念, 而是以顾客满意, 赢得顾客高忠诚度为经营理念。企业经营策略的焦点不再以争取或保持市场占有率为主, 而是以争取顾客满意为经营理念。因此, 营销策略的重心不再放在竞争对手身上而是放在顾客身上, 放在顾客现实的、潜在的需求上。当企业提供的产品和服务达到了顾客事先的期望值, 顾客就基本满意；如果远远超越顾客的期望值, 且远远高于其他同行, 顾客才真正满意；如果企业能不断地或长久地令顾客满意, 顾客就会忠诚。忠诚的顾客不仅会经常性地重复购买, 还会购买企业其它相关的产品或服务；忠诚的顾客不仅会积极向别人推荐他所买的产品, 而且对企业竞争者的促销活动具有免疫能力一个不满意的顾客会将不满意告诉16-20个人, 而每一个被告知者会再传播给12-15个人。这样, 一个不满意者会影响到二、三百人。在互联网普及的今天, 其影响则更大。据美国汽车业的调查, 一个满意者会引发8笔潜在的生意, 其中至少有一笔会成交。而另一项调查表明, 企业每增加5%的忠诚顾客, 利润就会增长25%-95%。一个企业的80%的利润来自20%的忠诚顾客；而获取一个新顾客的成本是维持一个老顾客成本的6倍。所以,美国著名学者唐?佩 珀斯指出: 决定一个企业成功与否的关键不是市场份额, 而是在于顾客份额。 于是, 企业纷纷通过广泛细致的市场调研、与消费者直接接触、顾客信息反馈等方式来了解顾客在各方面的现实需求和潜在需求。依靠对企业满意忠诚的销售、服务人员, 定期、定量地对顾客满意度进行综合测定, 以便准确地把握企业经营中与“顾客满意” 目标的差距及其重点领域, 从而进一步改善企业的经营活动。依靠高亲和力的企业文化、高效率的人文管理和全员共同努力, 不断地向顾客提供高附加值的产品, 高水准的亲情般的服
（ 二 〇 一 二 年 六 月 外文文献及翻译 题 目： About Buiding on the Structure Design 学生姓名： 学 院：土木工程学院 系 别：建筑工程系 专 业：土木工程(建筑工程方向) 班 级：土木08-4班 指导教师：
英文原文： Building construction concrete crack of prevention and processing Abstract The crack problem of concrete is a widespread existence but again difficult in solve of engineering actual problem, this text carried on a study analysis to a little bit familiar crack problem in the concrete engineering, and aim at concrete the circumstance put forward some prevention, processing measure. Keyword:Concrete crack prevention processing Foreword Concrete's ising 1 kind is anticipate by the freestone bone, cement, water and other mixture but formation of the in addition material of quality brittleness not and all material.Because the concrete construction transform with oneself, control etc. a series problem, harden model of in the concrete existence numerous tiny hole, spirit cave and tiny crack, is exactly because these beginning start blemish of existence just make the concrete present one some not and all the characteristic of quality.The tiny crack is a kind of harmless crack and accept concrete heavy, defend Shen and a little bit other use function not a creation to endanger.But after the concrete be subjected to lotus carry, difference in temperature etc. function, tiny crack would continuously of expand with connect, end formation we can see without the
文献出处: Dalman, M. Deniz, and Junhong Min. "Marketing Strategy for Unusual Brand Differentiation: Trivial Attribute Effect." International Journal of Marketing Studies 6.5 (2014): 63-72. 原文 Marketing Strategy for Unusual Brand Differentiation: Trivial Attribute Effect Dalman, M. Deniz & Junhong Min Abstract This research investigates that brand differentiation creating superior values can be achieved not only by adding meaningful attributes but also meaningless attributes, which is called "trivial attribute effect." Two studies provided empirical evidences as following; first, trivial attribute effect creates a strong brand differentiation even after subjects realize that trivial attribute has no value. Second, trivial attribute effect is more pronounced in hedonic service category compared to the utilitarian category. Last, the amount of willingness to pay is higher when trivial attribute is presented and evaluated in joint evaluation mode than separate evaluation mode. Finally, we conclude with discussion and provide suggestions for further research. Keywords: brand differentiation, evaluation mode, service industry, trivial attribute Introduction Problem Definition Perhaps the most important factor for new product success is to create the meaningful brand differentiation that provides customers with superior values beyond what the competitors can offer in the same industry (Porter, 1985). Not surprisingly, more than 50 percent of annual sales in consumer product industries including automobiles, biotechnology, computer software, and pharmaceuticals are attributed to such meaningful brand differentiation by including new or noble attributes (Schilling &Hill, 1998). However, the brand differentiation that increases consumer preference is not only by introducing meaningful attributes but also meaningless attributes. For
7 Rigid-Frame Structures A rigid-frame high-rise structure typically comprises parallel or orthogonally arranged bents consisting of columns and girders with moment resistant joints. Resistance to horizontal loading is provided by the bending resistance of the columns, girders, and joints. The continuity of the frame also contributes to resisting gravity loading, by reducing the moments in the girders. The advantages of a rigid frame are the simplicity and convenience of its rectangular form.Its unobstructed arrangement, clear of bracing members and structural walls, allows freedom internally for the layout and externally for the fenestration. Rig id frames are considered economical for buildings of up to' about 25 stories, above which their drift resistance is costly to control. If, however, a rigid frame is combined with shear walls or cores, the resulting structure is very much stiffer so that its height potential may extend up to 50 stories or more. A flat plate structure is very similar to a rigid frame, but with slabs replacing the girders As with a rigid frame, horizontal and vertical loadings are resisted in a flat plate structure by the flexural continuity between the vertical and horizontal components. As highly redundant structures, rigid frames are designed initially on the basis of approximate analyses, after which more rigorous analyses and checks can be made. The procedure may typically inc lude the following stages: 1. Estimation of gravity load forces in girders and columns by approximate method. 2. Preliminary estimate of member sizes based on gravity load forces with arbitrary increase in sizes to allow for horizontal loading. 3. Approximate allocation of horizontal loading to bents and preliminary analysis of member forces in bents. 4. Check on drift and adjustment of member sizes if necessary. 5. Check on strength of members for worst combination of gravity and horizontal loading, and adjustment of member sizes if necessary. 6. Computer analysis of total structure for more accurate check on member strengths and drift, with further adjustment of sizes where required. This stage may include the second-order P-Delta effects of gravity loading on the member forces and drift.. 7. Detailed design of members and connections.
学校 毕业设计（论文）附件 外文文献翻译 学号： xxxxx 姓名： xxx 所在系别： xxxxx 专业班级： xxx 指导教师： xxxx 原文标题： Building construction concrete crack of prevention and processing 2012年月日 .
建筑施工混凝土裂缝的预防与处理1 摘要 混凝土的裂缝问题是一个普遍存在而又难于解决的工程实际问题，本文对混凝土工程中常见的一些裂缝问题进行了探讨分析，并针对具体情况提出了一些预防、处理措施。 关键词：混凝土裂缝预防处理 前言 混凝土是一种由砂石骨料、水泥、水及其他外加材料混合而形成的非均质脆性材料。由于混凝土施工和本身变形、约束等一系列问题，硬化成型的混凝土中存在着众多的微孔隙、气穴和微裂缝，正是由于这些初始缺陷的存在才使混凝土呈现出一些非均质的特性。微裂缝通常是一种无害裂缝，对混凝土的承重、防渗及其他一些使用功能不产生危害。但是在混凝土受到荷载、温差等作用之后，微裂缝就会不断的扩展和连通，最终形成我们肉眼可见的宏观裂缝，也就是混凝土工程中常说的裂缝。 混凝土建筑和构件通常都是带缝工作的，由于裂缝的存在和发展通常会使内部的钢筋等材料产生腐蚀，降低钢筋混凝土材料的承载能力、耐久性及抗渗能力，影响建筑物的外观、使用寿命，严重者将会威胁到人们的生命和财产安全。很多工程的失事都是由于裂缝的不稳定发展所致。近代科学研究和大量的混凝土工程实践证明，在混凝土工程中裂缝问题是不可避免的，在一定的范围内也是可以接受的，只是要采取有效的措施将其危害程度控制在一定的范围之内。钢筋混凝土规范也明确规定：有些结构在所处的不同条件下，允许存在一定宽度的裂缝。但在施工中应尽量采取有效措施控制裂缝产生，使结构尽可能不出现裂缝或尽量减少裂缝的数量和宽度，尤其要尽量避免有害裂缝的出现，从而确保工程质量。 混凝土裂缝产生的原因很多，有变形引起的裂缝：如温度变化、收缩、膨胀、不均匀沉陷等原因引起的裂缝；有外载作用引起的裂缝；有养护环境不当和化学作用引起的裂缝等等。在实际工程中要区别对待，根据实际情况解决问题。 混凝土工程中常见裂缝及预防: 1.干缩裂缝及预防 干缩裂缝多出现在混凝土养护结束后的一段时间或是混凝土浇筑完毕后的一周左右。水泥浆中水分的蒸发会产生干缩，且这种收缩是不可逆的。干缩裂缝的产生主要是由于混凝土内外水分蒸发程度不同而导致变形不同的结果：混凝土受外部条件的影响，表面水分损失过快，变形较大，内部湿度变化较小变形较小，较大的表面干缩变形受到混凝土内部约束，产生较大拉应力而产生裂缝。相对湿度越低，水泥浆体干缩越大，干缩裂缝越易产 1原文出处及作者：《加拿大土木工程学报》